Engineering and Construction

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The most important property of a strut as far as the determination of its load- carrying capacity is concerned is its slenderness, l, defined as the ratio of its effective length, LE, divided by the appropriate radius of gyration, r. Codes of practice such as BS 5950 used to place upper limits on l, as indicated in Table 15.1, so as to avoid the use of flimsy construction, i.e. to ensure that a member which will ordinarily be subject only to axial load does have some limited resistance to an acci- dental lateral load, does not rattle, etc. Although not now explicitly stated in the 2000 version of the code, it is still good practice to aim for robust construction.


Similarly, for BS 5400: Part 3, no actual limits are specified but the user is provided with a general advisory note to the effect that construction should be suitably robust. By contrast, BS 5400 does place upper bounds on local plate slenderness to avoid consideration of local buckling (see Table 15.2).

Strut design will normally require that, once a trial member has been selected and its loading and support conditions determined, attention be given to whichever of the following checks are relevant for the particular application:

(1) Overall flexural buckling – largely controlled by the slenderness ratio, l, which is a function of member length, cross-sectional shape and the support condi- tions provided; also influenced by the type of member
(2) Local buckling – controlled by the width-to-thickness ratios of the component plate elements (see Chapter 13); with some care in the original  choice of member this need not involve any actual calculation
(3) Buckling of component parts – only relevant for built-up sections such as laced and battened columns; the strength of individual parts must be checked, often by simply limiting distances between points of interconnection
(4) Torsional or torsional-flexural buckling – for cold-formed sections and in extreme cases of unusually shaped heavier open sections, the inherent low  torsional stiffness of the member may make this form of buckling more critical than simple flexural buckling.

In principle, local buckling and overall buckling (flexural or torsional) should always be checked. In practice, provided cross sections that at least meet the semi-compact limits for pure compression are used, then no local buckling check is necessary since the cross section will be fully effective.

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Various types of steel section may be used as struts to resist  compressive loads; Fig. 15.1 illustrates a number of them. Practical considerations such as the methods to be employed for making connections often influence the choice, especially for light members. Although closed sections such as tubes are theoretically the most efficient, it is normally much easier to make simple site connections, using the minimum of skilled labour or special equipment, to open sections. Typical arrangements include:

(1) light trusses and bracing – angles (including compound angles back to back) and tees
(2) larger trusses – circular hollow sections, rectangular hollow sections, compound sections and universal columns
(3) frames – universal columns, fabricated sections e.g. reinforced UCs
(4) bridges – box columns
(5) power stations – stiffened box columns.

Fig. 15.1 Typical column cross sections

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Introduction
Members subject to compression, referred to as either ‘columns’ or ‘struts’, form one of the basic types of load-carrying component.They may be found, for example, as vertical columns in building frames, in the compression chords of a bridge truss or in any position in a space frame.

In many practical situations struts are not subject solely to compression but, depending upon the exact nature of the load path through the structure, are also required to resist some degree of bending. For example, a corner column in a building is normally bent about both axes by the action of the beam  loads, a strut in a space frame is not necessarily loaded concentrically, the compression chord of a roof truss may also be required to carry some lateral loads. Thus many compression members are actually designed for combined loading as beam-columns. Notwith-standing this, the ability to determine the compressive resistance of members is of fundamental importance in design, both for the struts loaded only in compression and as one component in the interaction type of approach normally used for beam- column design.

The most significant factor that must be considered in the design of struts is buckling. Depending on the type of member and the particular application under consideration, this may take several forms. One of these, local buckling of individual plate elements in compression, has already been considered in Chapter 13.Much of this chapter is devoted to the consideration of the way in which buckling is handled in strut design.

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Determination of member sizes
Although the engineer may use his experience or other methods to determine preliminary member sizes, the graphs in Figs 11.5, 11.6 and 11.7 will be used for this example. In order to use these graphs, four parameters are required:

Fig. 11.8 Portal frame design example

It is suggested that the next stage is to check the overall stability of the frame. The main reason for this is that the only way to correct insufficient stability is to change the main member sizes. If any other checks are not satisfied, additional bracing can frequently be used to rectify the situation, without altering the member sizes.

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The design process explained here assumes that the reader is familiar with the basic concepts of limit state steel design, and the basic analysis and design of continuous beams. These notes take the form of a worked example.A method of determining preliminary member sizes has been included.

1 Methods of analysis BS 5950: Part 1allows two main methods of analysis of a structure:

(1) Linear elastic. The frame is analysed either by hand or by computer assuming linear elastic behaviour. Once the forces, moments and shears have been derived by elastic analysis the ultimate capacity of each section is checked using the rules given in Section 4 of the Code.

(2) Simple plastic theory. The frame is analysed using the basic principles of simple plastic theory.Once the forces,moments and shears have been derived by analysis the member capacities are checked. Those containing plastic  hinges are checked in accordance with Section 4.

Elastic analysis
Using elastic analysis it is to be expected that the structure will be heavier than that designed by plastic methods, but less stability bracing will be needed. It may well be that the final details will also be more simple.

It will remain the engineer’s responsibility to ensure that stability is provided both locally and in the overall condition. BS 5950: Part 12 provides no specific rules regarding the stability of the frame as a whole; this means that the engineer must ensure that the stability is checked using the general rules for all frames. He must also check that the movement of the frame under all loading cases is not sufficient to cause damage to adjacent construction, i.e. brick walls or cladding, the serviceability limit state of deflection.

Plastic analysis
The method of calculating the ultimate load of a portal frame is described in many publications.The main essence of the method is to assume that plastic ‘hinges’ occur at points in the frame where the value of M/Mp is at its highest value, the load being considered as increasing proportionally until the failure or ultimate state is reached.

Because of the straining at the hinge points it is essential that the local buckling and lateral distortion do not occur before failure. Failure is deemed to have taken place when sufficient hinges have formed to create a mechanism.

The member capacities are calculated using the rules given in Section 4 of the Code but with the additional restrictions applied to hinge positions. In addition, positive requirements are put on checking frame stability for both single-bay and multi-bay frames. Plastic designed frames are lighter than elastic designed frames, providing deflection is not a governing point; however, additional bracing may well be required.

2 Stability
With the use of lighter frames, various aspects of stability take a more prominent part in the design procedures.As far as portal frames are concerned the following areas are important:

(1) overall frame stability, in that the strength of the frame should not be affected by changes in geometry during loading (PD effect)
(2) snap-through stability, in multi-bay frames (three or more), where the effects of continuity can result in slender rafters
(3) plastic hinge stability, where the member must be prevented from moving out of plane or rotating at plastic hinges
(4) rafter stability, ensuring that the rafter is stable in bending as an unrestrained beam
(5) leg stability, where the leg below the plastic hinge must be stable
(6) haunch stability, where the tapered member is checked to ensure that the inner (compression flange) is stable.

3 Selecting suitable members for a trial design
The design of a portal frame structure is in reality a process of selecting suitable members and then proving their ability to perform in a satisfactory manner. Inex- perienced engineers can be given some guidance to estimate initial member sizes.

In order to speed the initial selection of members, three graphs have been produced to enable simple pin-based frames to be sized quickly.

These graphs have been prepared making the following assumptions:

(1) plastic hinges are formed at the bottom of the haunch in the leg and near the apex in the rafter, the exact position being determined by the frame geometry
(2) the depth of the rafter is approximately span/55 and the depth of the haunch below the eaves intersection is 1.5 times rafter depth
(3) the haunch length is 10% of the span of the frame, a limit generally regarded as providing a balance between economy and stability
(4) the moment in the rafter at the top of the haunch is 0.87Mp, i.e. it is assumed that the haunch area remains elastic
(5) the calculations assume that the calculated values of Mp are provided exactly by the sections and that there are no stability problems. Clearly these conditions will not be met, and it is the engineer’s responsibility to ensure that the chosen sections are fully checked for all aspects of behaviour.

The graphs cover the range of span/eaves height between 2 and 5 and rise/span of 0 to 0.2 (where 0 is a flat roof). Interpolation is permissible but extrapolation is not.

The three graphs give:

Figure 11.5: the horizontal force at the feet of the frame as a proportion of the totalfactored load wL, where w is the load/unit length of rafter and L is the span of the frame
Figure 11.6: the value of the moment capacity required in the rafters as a proportion of the load times span wL^2
Figure 11.7: the value of the moment capacity required in the legs as a proportion of the load times span wL^2.

Fig. 11.5 Rise/span against horizontal force at base for various span/eaves heights

Fig. 11.6 Rise/span against required Mp of rafter for various span/eaves heights

Fig. 11.7 Rise/span against required Mp of leg for various span/eaves heights

The graphs are non-dimensional and may be used with any consistent set of units.

In the worked example kilonewtons and metres are used.

Method of use of the graphs

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General
The formulae given in this section are based on Professor Kleinlogel’s Rahmen-formeln  and Mehrstielige Rahmen.

The formulae are applicable to frames which are symmetrical about a central vertical axis, and in which each member has constant second moment of area.

Formulae are given for the following types of frame:

Frame I Hingeless rectangular portal frame.
Frame II Two-hinged rectangular portal frame.
Frame III Hingeless gable frame with vertical legs.
Frame IV Two-hinged gable frame with vertical legs.

The loadings are so arranged that dead, snow and wind loads may be reproduced on all the frames. For example, wind suction acting normal to the sloping rafters of a building may be divided into horizontal and vertical components, for which appropriate formulae are given, although all the signs must be reversed because the loadings shown in the tables act inwards, not outwards as in the case of suction.

It should be noted that, with few exceptions, the loads between node or panel points are uniformly distributed over the whole member. It is appreciated that it is normal practice to impose loads on frames through purlins, siderails or beams. By using the coefficients in Fig. 11.1, however, allowance can be made for many other symmetrically placed loads on the cross-beams of frames I and II shown, where the difference in effect is sufficient to warrant the corrections being made. The inde- terminate BMs in the whole frame are calculated as though the loads were uniformly distributed over the beam being considered, and then all  are adjusted by multiplying by the appropriate coefficient in Fig. 11.1. It may be of interest to state why these adjustments are made. In any statically indeterminate structure the inde- terminate moments vary directly with the value of the following quantity:


Where the loaded member is of constant cross section, EI may be ignored.

Fig. 11.1 Conversion coefficients for symmetrical loads

Consider, as an example, the case of an encastré beam of constant cross section and of length L carrying a UDL of W. Then the area of the free BM diagram is


If, however, W is a central point load, the area of the free BM diagram is

The fixed end moments (FEM) due to the two types of loadings are WL/12 and WL/8 respectively, thus demonstrating that the indeterminate moments vary with the area of the free BM diagram and proving that the indeterminate moments are in the proportion of 1 : 1.5.

No rules can be laid down for the effect on the reactions of a change in the mode of application of the load, although sometimes they will vary with the indeterminate moments. Consider a simple rectangular portal with hinged feet. If a UDL placed over the whole of the beam is replaced by a central point load of the same magnitude, then the knee moments will increase by 50% with a corresponding increase in the horizontal thrusts H, while the vertical reactions V will remain the same.

Although the foregoing remarks relating to the indeterminate moments resulting from symmetrical loads apply to all rectangular portals, the rule applies for asym- metrical loads imposed upon the cross-beam of a rectangular portal frame with hinged feet. If a vertical UDL on the cross-beam is replaced by any vertical load of the same magnitude, then the indeterminate moments vary with the areas of the respective free BM diagram.

No doubt readers who use the tables frequently will learn short cuts, but it is not inappropriate to mention some. For example, if a UDL of W over the whole of a single-bay symmetrical frame is replaced by a UDL of the same magnitude of W over either the left-hand or right-hand half of the frame, the horizontal thrust at the feet is unaltered. If the frame has a pitched roof then the ridge moment will also be unaltered.

The charts in the Appendix have been prepared to assist in the design of rectangular frames or frames with a roof pitch of 1 in 5.

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The solution of this type of beam consists, in the first instance, of the evaluation of the fixing or negative moments at the supports.

The most general method is the use of Clapeyron’s Theorem of Three Moments.

The theorem applies only to any two adjacent spans in a continuous beam and in its simplest form deals with a beam which has all the supports at the same level, and has a constant section throughout its length.

The proof of the theorem results in the following expression:

where MA,MB and MC are the numerical values of the hogging moments at the sup- ports A, B and C respectively, and the remaining terms are illustrated in Fig. 10.7.

In a continuous beam the conditions at the end supports are usually known, and these conditions provide starting points for the solution.

The types of end conditions are three in number:

(1) simply supported
(2) partially fixed, e.g. a cantilever
(3) completely fixed, i.e. the end of the beam is horizontal as in the case of a fixed beam.

The SF at the end of any span is calculated after the support moments have been evaluated, in the same manner as for a fixed beam, each span being treated  separately.

It is essential to note the difference between SF and reaction at any support, e.g. with reference to Fig. 10.7 the SF at support B due to span AB added to the SF at B due to span BC is equal to the total reaction at the support.

If the section of the beam is not constant over its whole length, but remains  constant for each span, the expression for the moments is rewritten as follows:


in which I1 is the second moment of area for span L1 and I2 is the second moment of area for span L2.

Fig. 10.7 Clapeyron’s Theorem of Three Moments

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Propped cantilevers
Beams which are built-in at one end and simply-supported at the other are known as propped cantilevers. Normally, the ends of the beams are on the same level, in which case bending moments and reactions may be derived in two ways: by employing the Theorem of Three Moments or by deflection formulae.Appropriate formulae for propped cantilevers under various types of loading are  presented in the Appendix Bending moment, shear and deflection tables for propped cantilevers.

Sinking of supports
When the supports for a loaded propped cantilever do not maintain the same relative levels as in the unloaded condition, the BM and SF may be obtained by using the deflection method (Fig. 10.2).When the prop, B, sinks the load which it takes is reduced, while the fixing moment at the other end is increased. Two special cases arise: the first when the prop sinks so much that no load is taken by the prop, and the second when the built-in end sinks so much that the fixing moment is reduced to zero, i.e. the cantilever resembles a simple support beam. The two special cases are shown in Fig. 10.2.

Fig. 10.2 Bending moment diagram for propped cantilever with sinking support

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The calculations required to obtain the shear forces (SF) and bending moments (BM) in simply-supported beams form the basis of many other calculations required for the analysis of built-in beams, continuous beams and other indeterminate  structures.

Appropriate formulae for simple beams and cantilevers under various types of loads are presented in the Appendix Bending moment, shear and deflection tables for cantilevers and simply-supported beams.

In the case of simple beams it is necessary to calculate the support reactions before the bending moments can be evaluated; the procedure is reversed for built- in or continuous beams. The following rules relate to the SF and BM diagrams for beams:

(1) the shear force at any section is the algebraic sum of normal forces acting to one side of the section
(2) shear is considered positive when the shear force calculated as above is upwards to the left of the section
(3) the BM at any section is the algebraic sum of the moments about that section of all forces to one side of the section
(4) moments are considered positive when the middle of a beam sags with respect to its ends or when tension occurs in the lower fibres of the beam
(5) for point loads only the SF diagram will consist of a series of horizontal and vertical lines,while the BM diagram will consist of sloping straight lines, changes of slope occurring only at the loads
(6) for uniformly distributed loads (UDL) the SF diagram will consist of sloping straight lines, while the BM diagram will consist of second-degree parabolas
(7) the maximum BM occurs at the point of zero shear, where such exists, or at the point where the shear force curve crosses the base line.

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The finite element idealization should represent the real structure as closely as possible with regard to geometrical shape, loading and boundary conditions. The geometrical form of the structure is the major factor to be considered when deciding the shape of elements to be used. In two-dimensional analyses the most frequently used elements are triangular or rectangular shapes. The triangular element has the advantage of simplicity in use and the ability to fit into irregular boundaries. Figure 9.29 shows an example using triangular elements and Fig. 9.30 shows a combination of triangular and rectangular elements. These figures also demonstrate the need to use relatively small elements in areas where high stress gradients occur. In many such cases the geometrical shape of the structure is such that  a  fine mesh of  elements is required in order to match this shape.

Fig. 9.29 Circular hole in uniform stress field

Fig. 9.30 Perforated tension strip (plane stress) – a quarter of a plate is analysed

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The evaluation of the stress resultants in members of skeletal frames involves the solution of a number of simultaneous equations.When a structure is in equilibrium, every element or constituent part of it is also in equilibrium. This property is made use of in developing the concept of the free body diagram for elements of a structure.

The portal frame sketched in Fig. 9.16 will now be considered for illustrating the concept. Assuming that there is an imaginary cut at E on the beam BC, the part ABE continues to be in equilibrium if the two forces and moment which existed at section E of the uncut frame are applied externally.The internal forces which existed at E are given by (1) an axial force F, (2) a shear force V and (3) a bending moment M. These are known as stress resultants. The external forces on ABE, together with the forces F, V and M, keep the part ABE in equilibrium; Fig. 9.16(b) is called the free body diagram. On a rigid jointed plane frame there are three stress resultants at each imaginary cut. The part ECD must also remain in equilibrium. This  consideration leads to a similar set of forces F, V and M shown in Fig. 9.16(c). It  will be noted that the forces acting on the cut face E are equal and opposite. If the two free body diagrams are moved towards each other, it is obvious the internal forces F, V and M cancel out and the structure is restored to its original state of equilibrium.As previously stated, equilibrium implies ΣPx = 0; ΣPy = 0; ΣM = 0 for a planar structure. These equations can be validly applied by considering the structure as a whole, or by considering the free body diagram of a part of a structure.

In a similar manner, it can be seen that a three-dimensional rigid-jointed  frame has six stress resultants across each section. These are the axial force, two shears in two mutually perpendicular directions and three moments, as shown in Fig. 9.17.

Fig. 9.16 Free body diagram

Fig. 9.17 Force and moments in x, y and z directions

With pin-jointed frames, be they two- or three-dimensional, there is only one stress resultant per member, viz. its axial load.When forces act on an elastic structure, it undergoes deformations, causing displacements at every point within the structure.

The solution of forces in the frames is accomplished by relating the stress result- ants to the displacements. The number of equations needed is governed by the degrees of freedom, i.e. the number of possible component displacements. At one end of the member of a pin-jointed plane frame, the member displacement has translational components in the x and y directions only, and no rotational displacement. The number of degrees of freedom is two. By similar reasoning it will be apparent that the number of degrees of freedom for a rigid-jointed plane frame member is three. For a member of a three-dimensional pin-jointed frame it is also three, and for a similar rigid-jointed frame it is six.

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Commercial buildings use energy primarily for space and water heating, lighting, cooling, ventilation and small power. In order to reduce operational energy, buildings should be designed to provide maximum occupant comfort throughout the year with minimum energy requirements. The charts in Fig. 8.2 show typical energy requirements and carbon dioxide emissions for two types of commercial building.

Fig. 8.2 Comparison of operational energy requirements and resulting carbon emissions for a typical air conditioned office and a good practice naturally ventilated open plan office (source: Econ 19, Energy use in offices, Energy Efficiency Best Practice Programme)

Although heating accounts for a significant part of the annual operational energy use, it produces a relatively small proportion of the overall carbon emissions since it is generally provided by burning natural gas. In the UK, gas produces only 41% of the carbon dioxide emissions per kilowatt hour that are produced by electricity.

Therefore the greatest scope for reducing carbon emissions from commercial buildings is in reducing the energy required for cooling, mechanical ventilation and artificial lighting.

Energy use is heavily dependent upon building design, including such factors as layout, orientation, thermal capacity, glazing arrangements, solar  shading, cooling and ventilation.The choice of structural system (steel or concrete frame4 ) has been found to have very little effect on operational energy requirements, so the follow- ing notes apply in general to all commercial buildings.

Structural steel gives the design team the flexibility it needs  to exploit these factors to the full.

One of the major ways of providing the necessary comfort levels is to make use of fabric energy storage.

This involves using the floor plate to absorb heat during the period when the building is occupied during the day and then purging it at night as shown in Fig. 8.3.

Initially there was a perception that thick concrete floors were needed for the system to be effective. However, it has been shown that, with a 24-hour cycle of heating and cooling, the necessary volume and disposition of concrete can be provided by composite construction.

Perforated permeable ceilings can be used to permit the necessary contact between the warm air and the floor. Enhancements can be achieved by ducting air immediately above or below the slab. Examples are illustrated in the SCI publication Environmental Floor Systems.

Fig. 8.3 Diurnal cycle for fabric energy storage

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Residential buildings account for about 30% of the total energy used in the UK. This relatively high percentage of energy use, together with increasing demand for new housing, indicates the importance of achieving improved energy efficiency in this sector.

Light steel frame residential buildings have proved to have good standards of environmental performance, particularly in terms of energy conservation. Many have significantly exceeded the requirements of UK Building Regulations in terms of thermal performance, as shown in Table 8.1.

Table 8.1 Comparison of thermal performance of light steel frame residential buildings with UK Building Regulations

The following measures are recommended to reduce operational energy  requirements:

• Minimization of heat loss through the building envelope
• Provision of an airtight building envelope with controlled ventilation
• Provision of efficient and controlled heating
• Use of appropriate passive heating, cooling and ventilation methods.

These measures can often be achieved at little or no extra overall cost, particularly where savings can be made in the capital cost of heating systems as a result of additional insulation and airtight construction. To accompany the regulations, a document is being developed showing robust details of plans which will, if properly constructed, give compliance with the UK Building Regulations. To achieve compliance with the new regulations will be easier with framed construction than  with traditional methods.

Light steel frame housing has several generic advantages but as with all forms of construction attention should be paid to the detailing in order to maximize the benefits.

Placing the insulation on the outside of the frame is known as warm frame construction; this minimizes heat loss since there is minimal thermal bridging. Risk of condensation is eliminated since all the members are contained within the warm internal environment. High levels of insulation are possible however without  significantly increasing wall thickness by the addition of insulation between the frame members. Heat loss resulting from air infiltration can also be conveniently minimized by the inclusion of an air tight membrane in the walls.

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The embodied energy burdens of commercial buildings vary considerably depending upon their specification, as illustrated in Fig. 8.1. For example, prestige office buildings can have twice the embodied energy content of lower specification offices.

This is due to the way they are clad and fitted out rather than  the effects of the structural frame.

Embodied energy is typically small in comparison to operational energy over the life of a commercial building (generally an order of magnitude less).The balance is however changing as improved designs reduce operational energy and the rate of change of use increases so requiring more refurbishment.

Fig. 8.1 Comparison of embedded energy and operating energy for a basic  grade and prestige office over a 60-year life

The embodied energy of the structural components of a typical office building is only about 25–30% of the total embodied energy and tends to vary relatively little irrespective of the type of structure.A life cycle assessment (LCA) study of a typical four storey office building compared five alternative steel and concrete structures.

No significant difference was identified between the embodied energy burdens of the different options.Although steel has a high embodied energy content per kilo- gram, its high strength to weight ratio means that a much lower volume of steel is required to construct a steel frame than the volume of concrete needed to construct an equivalent frame. The cross-sectional area and mass of steel structural members is very much smaller than that of equivalent concrete members.

The comparative lightness of modern steel frame structures has been important in reducing their embodied energy burdens. In addition to savings associated with the frame materials, designers can often adopt lighter foundation systems than would be required for other forms of construction. Alternatively the number of foundations may be reduced, since the spanning capabilities of steel can allow larger structural bays, and consequently fewer column bases requiring support.The larger bays also allow for more flexible use of space; steel frames are  easily adaptable, thereby prolonging the building life and reducing yet further the impact of embodied energy over the life cycle.

In composite construction, where beams act in tandem with the floor slab, the additional strength and stiffness that are achieved allow the use of relatively shallow beams. These have lower embodied energy and the volume of concrete in the floor is barely altered.

Modern portal frame design has played a significant role in reducing the em- bodied energy burden of single storey industrial and commercial buildings. Portal frames are capable of spanning considerable distances using relatively light sections.

Roof beam depths are generally of the order of span divided by 60, as opposed to span divided by 25 for a conventional beam and column roof. The combination of steel frames with light gauge purlins and rails supporting insulated cladding systems provides a very efficient form of construction resulting in the very high (95%) major share. Systems are available to meet the requirements of the new Part L to the UK Building Regulationsin terms of both insulation and air tightness.

A key element in achieving low embodied energy is the pursuit of structural  efficiency. By reducing the amount of structural steel required per unit area  of commercial building, designers can not only achieve elegant and economic  structural solutions, but also reduce embodied energy burdens.

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The UK steelwork sector has worked and continues to work hard to provide the most economic solutions and work methods. It has become world-class through its investments in IT and associated technologies, based on the use of 3-D modelling for detailing with direct links to CNC machines and suppliers. This is proving to be a strong foundation for the release of further benefits which stems from integration with design activities through the adoption of data transfer protocols such as the CIMSteel Integration Standards.

Since long before the needs of sustainability were fully recognized, there has been a history of continuous improvement in structural design methods which have resulted in economies to clients over a period of years.

This has been encouraged and has been complemented by the development of systems and components,many involving the use of light gauge steel in addition to the more traditional hot rolled sections.

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Introduction
The UK Government, along with many others, has a very strong policy to encour- age sustainable development. In May 1999, it published A better quality of life – a strategy for sustainable development for the United Kingdom. The principles in this document were further developed for the construction industry in Building a better quality of life – a strategy for more sustainable construction, which was published in April 2000.The underlying principles of sustainability lie in the appropriate balance of economic, social and environmental impacts, the so-called triple bottom line.The strong message is that for a viable long term future any enterprise must pay due concern to the following issues:

• Maintenance of high and stable levels of economic growth and employment
• Social progress which recognizes the needs of everyone
• Prudent use of natural resources
• Effective protection of the environment.

Clearly, to be successful all these issues need to be balanced, as over-commitment to any one of them at the expense of the others will lead to failure in the long term.

There are many examples of how the steel construction sector has progressed to be itself more sustainable and contribute to sustainable construction. Some of these are set out below.

Steel has a significant role in construction and has established a strong position.

In the UK structural frames now have approximately 70% of the multi-storey frame and 95% of the single storey industrial buildings market.

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The design requirements for steel structures in which brittle fracture is a consideration are given in most structural codes.There are several key factors which need to be considered when determining the risk of brittle fracture in a structure. These are:

(1) minimum operating temperature
(2) loading – in particular, rate of loading
(3) metallurgical features such as parent plate, weld metal or HAZ
(4) thickness of material to be used.

Each of these factors influences the likelihood of brittle fracture occurring.

From experimental data and parametric studies examining the maximum tolerable defect sizes in steels under various operating conditions, codes such as BS 5400-3 and BS 5950-1 provide tables giving the maximum permissible thickness of steel at given operating temperatures.Where a design does not fit into these broad categories attempts have been made to use fracture mechanics to provide a criterion for material selection in terms of Charpy test energy absorption as this information is usually provided by the steel makers. When true fracture mechanics toughness values are available, methods such as those described in References 8 and 9 should be employed to assess the acceptability of flaws in fusion welded structures.

As discussed earlier, at normal operating temperatures and slow rates of loading valid KIc values are not usually obtained for structural steels, and in the offshore and nuclear industries critical CTOD and J tests are widely used. Under these conditions, valid KIc values from low temperature tests will be conservative. If the structure is acceptable when assessed using such conservative data then there is often no great need to pursue the problem, particularly as all forms of fracture mechanics testing are expensive compared with routine quality control tests such as Charpy testing.

In general, the fracture toughness of structural steel increases with increasing temperature and decreasing loading rates. The effect of temperature on fracture toughness is well known. The effect of loading rate may be equally important, not only in designing new structures, but also in understanding the behaviour of existing ones which may have been built from material with low toughness at their service temperature.The shift in the ductile–brittle transition temperature for structural steels can be considerable when comparing loading rates used in slow bend tests with those in Charpy tests. Results from experimental work have shown that the transition temperature for a BS EN 10025 S355 J steel can change from around 0°C to -60°C with decreasing loading rate.

Materials standards set limits for the transition temperatures  of various steel grades based on Charpy tests.When selecting an appropriate steel for a given structure it must be remembered that the Charpy values noted in the standards apply to parent plate. Material toughness varies in the weld and heat-affected zones of welded joints, and these should be checked for adequate toughness. Furthermore, since larger defects may be present in the weld area than the parent plate, appropriate procedures should be adopted to ensure that a welded structure will perform as designed. These could involve non-destructive testing including visual examination. In situations where defects are found, fracture mechanics procedures such as those in References 8 and 9 can be used to assess their significance.

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The structural implications of the above can be summarized as follows:

• Structural fire resistance: The required fire resistance for elements of structure is equivalent to that of the building without the atrium.
• Facade: The structure needs to ensure that the facade can achieve its  required performance.This may equate to ensuring that the deflections in a fire condition are limited or, more simply, by specifying a fire-resisting system for the facade.

Although BS 5588: Part 7 defines a performance criterion for ‘smoke retarding’ construction, consideration should be given to specifying fire-resisting construction instead; this is because there is no standard specification for smoke-retarding construction, and there will be necessity to develop ad-hoc material specification, construction methods and QA procedures to ensure that the system meets the required performance and to demonstrate this to the regulators.

• Atrium roof: Whilst structural elements solely supporting the roof neither  formally nor usually require fire resistance, it is prudent to consider the effect of collapse in development of the fire strategy. Thus potentially there may be a requirement for fire resistance. Also, as normal, where the roof structural elements support elements requiring a fire-resistance (such as the atrium facade), then these structural elements must achieve the required fire performance.

The fire strategy for a building with an atrium is complex, particularly where there is a need to incorporate local government Acts or property protection requirements, or where there are non-standard client, architectural, or engineering objectives.

A fire-engineered approach will enable the optimum strategy and best value to be achieved. Structural fire engineering can be used with an atrium building in the same way as for a building without an atrium, taking account of  the inherent  performance of the structure to reduce or omit the need for fire protection. Fire engineering enables best integration of the fire strategy within the design, to ensure that all systems are tailored to meet the precise performance and reliability requirements (for fire and all other requirements).

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Background
The fundamental principle of atrium fire safety design is to ensure that the design standard is not compromised by the presence of the atrium; the standard of the building with the atrium should not be less than the standard of an equivalent building without the atrium. This section describes the considerations and processes involved in the development of the atrium fire strategy, concluded by summarizing the implications for the structural design.

The primary objective of the fire strategy is to achieve an acceptable standard of life safety. BS 5588: Part 7 is the relevant ‘life-safety’ guidance for the incorporation of atria into buildings, and is applicable in all but limited circumstances.A package of fire safety provisions is developed to address the following requirements:

• Means of escape
• Compartmentation
• Structural fire resistance
• External spread of fire
• Facilities for the fire service. Smoke clearance is relevant for this situation.

This package depends on the size and use of the building.As examples:

(1) In a small office building, the design is for simultaneous evacuation and there is no need for compartmentation between storeys. Thus in this situation the atrium can be open, as the consequences of a fire anywhere in the building with an atrium are no different (in terms of life safety) than those in the same building without an atrium.
(2) In tall office buildings, storey-to-storey compartmentation is defined to enable phased evacuation; this vertical compartmentation needs to be maintained in the building with an atrium.
(3) In a hospital, as well as vertical compartmentation there is a need for horizontal compartmentation.This enables progressive horizontal evacuation of the fire compartment, with all other compartments remaining in place.
(4) With an atrium it is necessary to consider thermal radiation across the atrium, to ensure that this does not create untenable conditions in a non-fire compartment on the same storey adjacent to the atrium.

In situations where occupants remain in the building, the psychological effects of smoke in the atrium must also be considered, even though there may be no other risk posed by the fire.

The property protection and business continuity implications must be carefully considered. Even in situations where there is an acceptable standard of life safety, the atrium provides the opportunity for substantial spread of smoke, especially  with an unenclosed atrium. These additional requirements are usually specified by the client, operator, or insurer and not regulated under legislation. However, it should be noted that there are Local Government Acts (such as Section 20 in London) and national legislation outside England which have property protection implications.

The design procedure of the atrium can be generalized by consideration of:

• Use of the accommodation adjacent to the atrium
• Use of the atrium base
• Performance of the atrium façade
• Control of smoke within the atrium.

Effective smoke control is usually the key to the success of the atrium fire strategy.

This can be achieved either by preventing smoke entering the atrium (e.g. an atrium façade with fire-resisting performance and a fire-sterile atrium  base) or, alternatively, by managing any smoke within the atrium. The atrium can be divided into three zones:

(1) Atrium base
(2) The smoke reservoir at the top of the atrium, where any hot smoke will collect and can be extracted
(3) Remainder of the atrium.

Different design considerations will apply for these different zones.

The smoke management will normally take advantage of a smoke extract system.

The design of this system depends on the quantity of smoke within the atrium and the extract capacity. The quantity of smoke is dependent on the fire size and width of spill, where the smoke spills into the atrium. Complex spill geometries create large quantities of smoke. Smoke extract can be by either mechanical extract or natural systems. In either case it is necessary to arrange for low-level make-up for replacement air.Wind-effects on natural systems must be addressed.The extract can serve either to limit the depth of the smoke reservoir, or to limit the smoke reser- voir temperature, or both. In either case the fire size must be controlled.

The fire size is normally controlled either by use of sprinklers or by management control. Management control defines islands of fire load, the size  of the island dependent on the amount of combustibles, separated to ensure that the fire does not spread between islands. Typical rules are 10m2 islands separated by 4m.

Sprinklers are usually assumed in smoke control calculations to limit the fire size to a maximum area equal to the sprinkler-grid area. In practice, sprinklers may extinguish the fire or control it to a substantially smaller area.As an alternative to sprinklers or management control, a fire-engineered approach can  be used to account for the likelihood of a particular fire size occurring, given the use of the accommodation (‘natural fire’ based design).

It is important also to account for cool smoke, which cannot be assumed to rise (and may even drop).Where the smoke cannot be assumed to be hot enough to rise to the reservoir by buoyancy, it will be carried by the airflows that exist within the atrium. CFD analysis may be required to design the cold smoke system. Careful coordination with the building services engineer will enable the best value solution, particularly with natural ventilation systems, as the cold smoke is primarily carried by the environmental flows. An upward flow of air is beneficial for both the fire strategy and the environmental strategy.

Use of the atrium base can be enabled, provided that the rules for smoke and fire spread are respected. It is difficult to use sprinklers for controlling the fire size in the atrium base, as there is no ceiling above it (the atrium roof is usually too high to enable effective operation of sprinklers), and thus fire load control is either by management control or by an engineered approach.

Where the atrium facade is required to have a fire performance, this is defined either to prevent the fire breaking into the atrium, or to prevent the smoke breaking back in.Where an enclosure is required to have a fire-resisting performance, this can either be achieved by a formal fire-resisting system, or alternatively a smoke-retarding construction may be used in combination with an extract system to limit smoke temperatures. There is currently no standard for smoke-retarding construction, and thus it is often preferable to specify fire-resisting construction to avoid complicated specifications and enhanced site control.

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There are several factors that affect the cost of steel structures of this type:

• Type of section used
• Complexity of fabrication
• Complexity of installation
• Type of finish
• Quality of workmanship
• Procurement arrangements

It is common to budget steelwork on the basis of rates applied  to each tonne of material.While this is a useful indication of cost, it is important to recognize that the appropriate rates for these types of structure may be two to four times greater than those for standard structural steelwork.