Engineering and Construction

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Pad foundations are used primarily to support the major structural elements in either sheds or multi-storey buildings. The pad foundations to major elements may be either mass concrete or reinforced concrete, the latter when either heavy loads or very poor ground conditions are present. They may be used in  the context of cladding to support intermediate posts carrying sheeting rails, in which case the load is almost all from wind forces and is horizontal.

Strip foundations are used in steel-framed buildings to support external masonry or brickwork cladding and masonry internal partitions. In some cases the ground floor is thickened at these locations to provide a foundation but care should be taken with respect to the appropriate depth for clay or frost heave and for compatibility between such foundations and those of the main frame.

Piled foundations, either driven, bored or cast in place, are used on sites where ground conditions are poor or for buildings or structures in which differential set- tlement is critical. They may also be required in circumstances where heavy con- centrations of load occur. In general when piled foundations are used the whole of  the construction should be supported on piles. The ground floor slab, ground floor cladding and internal partitions should be carried by ground beams between the pile cap locations. If it is necessary for reasons of economy to support the ground floor independently, provision should be made for differential settlement by the inclusion of suitable movement joints.

Ground improvement techniques are appropriate for some types of poor ground.
 
The most usual techniques are vibro-compaction or vibro-replacement but dynamic
compaction can also be useful for improvement of large isolated sites. Ground improvement specialists or specialist consultants should be approached as economy will be the most important factor in the decision.
 
Typical foundation layouts are shown in Fig. 27.1.

Fig. 27.1 Part plan of typical two-bay crane shed.

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Welding offers many advantages over other joining methods:

• Freedom of design, and the opportunity to develop innovative structures
• Easy introduction of stiffening elements
• Less weight than in bolted joints because fewer plates are required
• Welded joints allow increased usable space in a structure
• Protection against the effects of fire and corrosion are easier and more effective.

1 Aesthetics and freedom of design
Probably the main benefit of welded construction is the freedom  of design, compared with bolted joints.

Some important types of structure, such as orthotropic bridge decks, Vierendeel trusses, tubular frames, tapered beams, and even most T-joints, could not be made as easily by any other method. Welded joints allow  more freedom in the use of rolled shapes, high strength steels, and corrosion- resistant steels.

Provided that the joints are made using appropriate materials and practices, designers should feel free to develop structures that are aesthetic as  well as functional.

Hollow sections, both rectangular and circular, can sensibly be joined only by welding, even if the final connection is by bolting.

2 Stiffness
Welded construction allows a designer to introduce stiffness and strength where required, in the most discreet and/or the most structurally efficient manner. The transfer of load and the stiffness can, with welding, be introduced in a gradual continuous manner, instead of in step changes through bolted pieces.

3 Weight, volume and size
Bolting requires the use of lap joints, which add weight to a structure. For example, the use of site welds, instead of bolted joints, to join plate girders to crossheads saved about 50 tonnes of connection plates, bolts, and nuts in the construction of the new Thelwall Viaduct.

In addition to savings in weight, the use of welded joints provides more usable volume in a typical building. Bolted joints reduce usable volume significantly because current legislation and fire engineering can demand that the fireproofing coatings that must be applied give a minimum thickness over every protruding nut and bolt head.

4 Durability and corrosion resistance
Total life cycle costing, including the costs of maintenance, is becoming increasingly important in all structures. For one aspect of this, the durability, the corrosion of structural steelwork can be retarded by three means – controlling the surrounding environment, effective coating systems, and the use of steels (and weld metals) with improved inherent corrosion resistance.Coating systems, however effective they are on plate surfaces, can only with great difficulty prevent corrosion in all the crevices of bolted joints, but the clean lines of welded structures allow the full effectiveness of modern coatings to be demonstrated.The use of corrosion resistant (weathering) steels (from BS EN 10155: 19931) imparts long life, and although they can be used in bolted connections, they perform poorly in crevices and thus they work far better in welded solutions.

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The traditional method of designing cased columns, presented in BS 449: 1969, can be very conservative but is readily accepted as a method of simple design.

The minimum width of the concrete casing, according to BS 449, should be the flange width b of the steel section plus 100mm.At least four reinforcing bars are to be located in the concrete at a cover not exceeding 50mm, to which 5mm diameter steel stirrups are attached at spacings not exceeding 200mm.The normal aggregate size is 10mm and the minimum concrete grade is 21N/mm2.

To establish the axial capacity of the cased column (or strut), the radius of gyration of the solid section is taken as 0.2 (b + 100) mm, or alternatively that of the major axis of the steel section. The cross section excludes any concrete cover  in excess of 75mm from the overall dimension of the steel section. The net area of concrete is replaced by an equivalent area of steel by dividing by a modular ratio of 30.

Therefore, knowing the permissible axial stress, as a function of the effective slen- derness of the cased column, and the equivalent area of steel, the axial capacity of the section can be easily calculated. However, an onerous limit on the use of this method is that axial load on the cased column should not exceed twice that permitted on the uncased section.

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The axial resistance of a stocky column, defined as a column that is not subject to the effects of instability, is determined by adding the ultimate compressive strengths of the steel and concrete components. In traditional reinforced concrete design, a ‘short’ column is defined as one whose effective height to least cross-sectional dimension is less than 12, and a nominal allowance is made for eccentricity of axial load.

The axial resistance of concrete-filled sections is greater because the concrete is not able to expand laterally (Poisson’s ratio effect) under load, and triaxial stresses are developed in the concrete. This causes an increase in the compressive strength of the concrete by an amount dependent on the proportions of the cross-section.

The hoop tensions created in the steel have a small adverse affect on its strength.

The effect of eccentricity of axial load is to develop a bending moment in the section. The moment resistance of the section (in the absence of axial load) can be calculated considering plastic stress blocks (see Fig. 22.2).

Fig. 22.2 Plastic stress blocks acting on cross-section of composite column subject to moment and axial force

The interaction between axial load and bending moment can be considered in terms of a simplified interaction formula.

Slender columns require a more refined treatment. The effective slenderness of a column is determined from the proportions of the composite section. The second moment of area is obtained by adding the second moments of area of the steel and concrete (divided by an appropriate modular ratio). This represents a considerable increase over the properties of steel alone.The axial stress that the section can resist is then determined from the column buckling curves for the steel section under consideration.The resulting axial stress, relative to the yield stress of the steel, is effectively a resistance reduction factor to be applied to the ‘stocky column’ compressive resistance.

It should be noted that the above approach assumes that loads are not applied laterally over the column length. Concentric and eccentric axial loads cause relatively low interface shear stresses between the steel and the concrete. Lateral loads cause greater shear stresses and may necessitate the introduction of shear-connectors.

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The traditional method of designing cased columns, presented in BS 449: 1969, can be very conservative but is readily accepted as a method of simple design.

The minimum width of the concrete casing, according to BS 449, should be the flange width b of the steel section plus 100mm.At least four reinforcing bars are to be located in the concrete at a cover not exceeding 50mm, to which 5mm diameter steel stirrups are attached at spacings not exceeding 200mm.The normal aggregate size is 10mm and the minimum concrete grade is 21N/mm2
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To establish the axial capacity of the cased column (or strut), the radius of  gyration of the solid section is taken as 0.2 (b + 100) mm, or alternatively that of the major axis of the steel section. The cross section excludes any concrete cover  in excess of 75mm from the overall dimension of the steel section. The net area of concrete is replaced by an equivalent area of steel by dividing by a modular ratio of 30.

Therefore, knowing the permissible axial stress, as a function of the effective slen- derness of the cased column, and the equivalent area of steel, the axial capacity of the section can be easily calculated. However, an onerous limit on the use of this method is that axial load on the cased column should not exceed twice that permitted on the uncased section.

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In many cases of design, concrete or some other protective material is required around steel columns for reasons of fire-resistance and durability. It would seem appropriate, therefore, to develop composite action between the steel and concrete, thereby taking advantage of the inherent compressive strength of the concrete, increasing the compressive resistance of the section and leading to considerable savings in the cost of the steelwork. Even ignoring this composite action, the slenderness of the steel column in lateral buckling is reduced, thereby increasing the compressive stress that can be resisted by the steel section.

Much research has gone into the behaviour of concrete-filled tubular sections.

Architecturally, tubular columns have many attractive features; concrete filling has no visual effect on their appeal.The advantages from a structural point of view are, first, the triaxial confinement of the concrete within the section, and second, the fire resistance of the column, which largely depends on the residual capacity of the concrete core.

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Composite columns comprise steel sections with a concrete encasement or core.

Encased columns usually consist of standard I-beam or H-column sections with a rectangular or square concrete section encasement to form a solid composite section. Additional reinforcement is placed in the concrete cover around the steel section in order to prevent spalling under axial stress and in fire.

Concrete-filled columns consist of circular, square or rectangular hollow sections filled with concrete.

Additional reinforcement is not normally required, except in columns of large section. Examples of these different types of composite columns are shown in Fig. 22.1.

The steel structure is normally constructed in advance of the formation of the composite section.The minimum size of the steel element is therefore controlled by the construction condition. For practical reasons it is normally necessary to fill the tubular sections shortly after their erection but I-sections can be encased much later.

However, there are practical problems in concreting around a column with the floors in place. Alternatively, columns can be pre-encased in the factory and in situ concrete used to fill the zone around the beam–column connections.

Fig. 22.1 Examples of composite columns: (a) RHS and CHS concrete-filled section, (b) concrete-encased section

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The modern form of shear-connector is the welded headed stud ranging in diameter from 13 to 25mm and from 65 to 125mm in height.The most popular size is 19mm diameter and 100mm height before welding.When used with steel decking, studs are often welded through the decking using a hand tool connected via a control unit to a power generator. Each stud takes only a few seconds to weld in place. Alternatively, the studs can be welded directly to the steel beams in the factory and the decking butted up to or slotted over the studs.

There are, however, some limitations to through-deck welding: the top flange of the beam must not be painted, the galvanized steel should be less than around  1.25mm thick, the deck should be clean and free of moisture, and there should be no gap between the underside of the decking and the top of the beam.The minimum flange thickness must not be less than the diameter of the stud  divided by 2.5  (typically, 19/2.5 = 7.6mm). The power generator needs 415V electrical supply, and the maximum cable length between the weld gun and the power control units should be limited to around 70m to avoid loss of power. Currently, only 13, 16 or 19mm diameter studs can be through-deck welded on site.

Where precast concrete planks are used, the positions of the shear-connectors are usually such that they project through holes in the slab which are later filled with concrete.Alternatively, a gap is left between the ends of the units sitting on the top flange of the beam on to which the shear connectors are fixed. Reinforcement (usually in the form of looped bars) is provided around the shear-connectors.

There is a range of other forms of welded shear-connector, but most lack practical applications. The ‘bar and hoop’ and ‘channel’ welded shear-connectors have been use in bridge construction.

Shot-fired shear-connectors may be used in smaller building projects where site power might be a problem.All shear-connectors should be capable of resisting uplift forces; hence the use of headed rather than plain studs.

The number of shear-connectors placed along the beam is usually sufficient to develop the full flexural resistance of the member. However it is possible to reduce the number of shear-connectors in cases where the moment resistance exceeds the applied moment and the shear-connectors have adequate ductility (or deformation capacity).

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In buildings and bridges, steel beams often support concrete slabs.Under load each component acts independently with relative movement or slip occurring at the interface. If the components are connected so that slip is eliminated, or considerably reduced, then the slab and steel beam act together as a composite unit (Fig. 21.1).

There is a consequent increase in the strength and stiffness of the composite beam relative to the sum of the components.

Fig. 21.1 Behaviour of composite and non-composite beams

The slab may be solid in situ concrete or the composite deck slab considered later. It may also comprise precast concrete units with an  in situ concrete topping. In buildings, steel beams are usually of standard UB scction, but UC and asymmetric beam sections are sometimes used where there is need to minimize the beam depth.A typical building under construction is shown in Fig. 21.2.Welded fabricated sections are often used for long-span beams in buildings and bridges.

Fig. 21.2 Composite building under construction showing decking and shear-connectors

Design of composite beams in buildings is now covered by BS 5950: Part 3, although guidance was formerly available in an SCI publication.

The design of composite beams incorporating composite slabs is affected by the shape and orientation of the decking, as indicated in Fig. 21.3.

Fig. 21.3 Composite beams incorporating composite deck slabs: (a) deck perpendicular to beam, (b) deck parallel to beam

One of the advantages of composite construction is smaller construction depths.
Services can usually be passed beneath, but there are circumstances where the beam depth is such that services can be passed through the structure, either by forming large openings, or by special design of the structural system.A good example of this is the stub-girder.

The bottom chord is a steel section and the upper chord is the concrete slab. Short steel sections or ‘stubs’ are introduced to transfer the forces between the chords.

Openings through the beam webs can be provided for services. Typically, these can be up to 70% of the beam depth and can be rectangular or circular in shape.

Examples of the above methods of introducing services within the structure are shown in Fig. 21.4.

Fig. 21.4 Different methods of incorporating services within the structural depth

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The decking supports the weight of concrete in the finished slab, excess weight from concrete placement and ponding, the weight of operatives and any impact loads during construction.There is a variety of recommendations for this temporary con- struction load to be considered in the design of the decking: these are expressed in terms of either a uniformly-distributed load or a single line load transverse to the span being checked. In BS 5950: Part 4, the standard used in the design of the decking and composite slab, these construction loads are specified as 1.5kN/m2 or a transverse line load of 2kN/m2  (for spans up to 3m) respectively. Self-weight loads are multiplied by a load factor of 1.41 and construction loads by 1.6.

However, significant loads can be developed before concreting due to storage  of equipment and materials.

Similarly, loads applied to the composite slab before it has gained adequate strength for fully composite action should not exceed  1.5kN/m2. The definition of ‘adequate strength’ is considered to be a cube strength of 75% of the specified value, which is often achieved in 5–7 days after concreting.

The design of single-span decking is usually controlled by deflection and ponding of concrete. Continuous decking is generally designed on strength rather than deflection criteria, and longer unpropped spans can be achieved. The construction load is applied in design as a pattern load even though only one span is likely to be loaded to this extent, in view of the progressive nature of concreting. The 1994 version of BS 5950: Part 4 reduced the construction load to 0.5kN/m2 on deck spans adjacent to the fully loaded one. In BS 59501
a limit on the residual deflection of the soffit of the slab after concreting of the smaller of (span/180) or (slab depth/10) is specified, increased to (span/130) if the effects of ponding are included. Greater deflections are likely to be experienced during concreting.

The design of the decking in bending is dependent on the properties of the profile, and particularly the thin plate elements in compression.Where profiles are stiffened by one or two folds in the compression plate, the section is more efficient in bending.

The design of continuous decking is based, according to code requirements, on an elastic distribution of moment, as a safe lower bound to the collapse strength. Elastic moments are normally greatest at internal supports. For the two equal span case this negative (hogging) moment is equal to that of the positive (sagging) moment of single-span decking.Many profiles with wide troughs are weaker under negative moment than positive, and the effect of the localized reaction at the internal support is to reduce further the negative moment resistance. In design to code requirements, therefore, continuous decking may appear to be weaker than simply- supported decking even though in reality it must be considerably stronger.

In general the load capacity of continuous decking is obtained  from tests. The reduced construction load on adjacent spans has the effect of reducing the design negative moment so that elastic design more closely approximates the results of tests under the full design load (see following section).

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Slab depths are normally in the range of 110–150mm for shallow profiles. Clearly, the minimum depth for structural adequacy is usually selected, but the slab depth is affected by insulation requirements in a fire (see Section 20.6.3) and serviceability criteria as influenced by the ratio of slab span to depth. As a general rule, span:depth ratios (considering the overall, slab depth) of continuous composite slabs should be less than 30 for LWC and 35 for NWC. Longer spans may suffer from excessive deflections and vibration of the supporting beams.

In designing the composite slab it is difficult to develop the full flexural resistance of the slab determined by the area of the deck acting as effective tensile reinforcement. Failure is normally one of shear-bond rupture resulting in slip between the deck and the concrete.Nevertheless, load capacities of composite slabs with the above proportions are normally adequate for most imposed loads up to 10kN/m2.The shear-bond resistance of composite slabs is determined from tests; this resistance may be enhanced by end-anchorage provided by shear-connectors attached to the support beams.

The concrete slab also acts as the principal compressive element in the design of the composite beams.This interaction between the flexural behaviour of the slab and the beams is only important where both the slab and beam span in the same direction.

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One of the principal advantages of steel decking is that it acts as a working  platform. The loads that it supports arise mainly during the concreting operation.

Lightweight concrete (LWC), therefore, has advantages in terms of its reduced self-weight in comparison to normal-weight concrete (NWC). In the UK the main form of LWC comprises Lytag and sand with a dry density range of 1800–1900kg/m3.The compressive strength of Lytag LWC is similar to NWC, although the elastic modulus is lower.

Concrete is usually placed by pump,mainly for reasons of speed, but also because it is difficult to ‘skip’ the concrete with the decking in place above. Indeed, it is usual practice to deck-out a number of floors ahead of the concreting  operation. The decking is attached to its supports by shot-fired pins or self-tapping screws and later by welded shear-connectors, welding or bolting.

Unlike traditional reinforced concrete, there is no need to restrict bay sizes during construction because the decking serves to distribute early age and shrinkage strains, thereby eliminating the formation of wide cracks.Typical pumped pours are 500–1000m2. LWC also has a higher tensile strain capacity than NWC, reducing the tendency for cracking.The mesh reinforcement is of a nominal size to control cracking at supports, and to act as ‘fire reinforcement’. A142 or A193 are the common sizes specified.

The concrete grade (cube strength in N/mm2) is normally specified as C30 to C40. Pumped concrete also contains additives to aid lubrication. For  this reason the slump test is not a good measure of workability and so the flow-meter method is often used. The concrete is usually tamped level by fixing the tamping rails to the support beams. As these beams deflect, the slab level adopts that  of the support beams. In propped construction, the slab level deflects further on removal of the props.

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Deck profiles are usually in the range of 38–80mm height and 150–300mm trough spacing with sheet thicknesses between 0.8mm and 1.5mm, use of the lower thickness being limited by local buckling and the upper thickness by difficulties in rolling.

A summary of the different decking profiles marketed for use in composite slabs is shown in Fig. 20.2. There are two well-known generic types: the dovetail profile and the trapezoidal profile with web indentations.

Fig. 20.2 Examples of modern composite deck profiles

The shape of the profile is controlled by a number of criteria:


(1) the need to maximize the efficiency of the cross-section in bending (both positive and negative moments)

(2) the need to develop adequate composite action with the concrete by use of embossments or indentations or by the shape of the profile itself

(3) the efficient transfer of shear from the beam into the concrete slab (a similar problem to the haunch design in reinforced concrete slabs).

Practice in North America has been to use profiles with a trough pitch typically  300mm and 50mm height to achieve a 3m span without temporary propping.

In the UK, spans of 3.5–3.75m are achieved with profiles of typically 60mm depth.

Slimflor ® construction2 has also led to the development of deep deck profiles of 210–225mm depth and 1.0–1.25mm thickness. The total slab depth of 280–310mm is such that the steel beam is fully encased in the slab except for its bottom flange.

The deep deck slab is designed for spans of 6m when unpropped during construction, or up to 7.5m when propped. The beam may take the form of a universal column section with a welded plate or three welded plates. Recent developments by Corus have resulted in the rolling of asymmetric beams.

These are rolled with a patterned top flange which enhances the composite action between the beam and
the overlying in situ concrete thus eliminating the need for welded shear connectors. The rolled sections have webs of greater thickness than that of the flanges to achieve superior fire resistance properties, and this extra web thickness also enhances the torsional properties of the section.

The grades of steel used for decking are specified in BS EN 10147. The common grades in the UK are S280 and S350. Steel decking is usually galvanized to a standard of 275 g/m2 (roughly 0.04mm per face), which gives adequate protection for internal use.

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Composite floor is the general term used to denote the composite action of steel beams and concrete or composite slabs that form a structural floor. Composite deck slabs, in this context, comprise profiled steel decking (or sheeting) as the permanent formwork to the underside of a concrete slab spanning between support beams. The decking acts compositely with the concrete under service loading.

It also supports the loads applied to it before the concrete has gained adequate strength. A light mesh reinforcement is placed in the concrete. A cross-section through a typical composite slab is shown in Fig. 20.1. Shear-connectors are used to develop composite action between the concrete slab and steel beams.

Fig. 20.1 Section through a typical composite deck slab

The decking has a number of roles. It

(1) supports the loads during construction
(2) acts as a working platform
(3) develops adequate composite action with the concrete
(4) transfers in-plane loads by diaphragm action to vertical bracing or walls (in conjunction with the concrete topping)
(5) stabilizes the beams against lateral buckling provided that the ribs run perpendicular to the beam or are at an angle of at least 45° to the beam
(6) acts as transverse reinforcement to the composite beams
(7) distributes shrinkage strains preventing serious cracking.

In addition it has a number of advantages over precast or in situ concrete alternatives:

(1) construction periods are reduced
(2) decking is easily handled
(3) attachments (e.g. ceiling hangers) can be made easily
(4) openings can be formed
(5) shear-connectors can be welded through the decking
(6) decking can be cut to length and is less prone to tolerance problems.

The main economy sought in buildings is speed of construction and for this reason slabs and beams are generally designed to be unpropped during the construction stage. Spans of the order of 3m to 3.6m between support beams are most common, but can be increased to over 4m if propping is used. Design of composite slabs is covered by BS 5950: Part 4.

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1 Use of Vierendeel girders
Vierendeel girders, unlike trusses or lattice girders, are rigidly-jointed open-web girders having only vertical members between the top and bottom chords. The chords are normally parallel or near parallel; some typical forms are shown in  Fig. 19.8(a).

The elements in Vierendeel girders are subjected to bending, axial and shear stress, unlike conventional trusses with diagonal web members where the members are primarily designed for axial loads. Vierendeel girders are usually more expensive than conventional trusses and their use is limited to instances where diagonal web members are either obtrusive or undesirable.Vierendeel girders in bridges are rare; they are more commonly used in buildings where access for circulation or a large number of services is required within the depth of the girder.

The economic proportions and span lengths are similar to those of the parallel chord trusses already discussed in section 19.2.

2 Analysis
Vierendeel girders are statically indeterminate structures but  various manual methods of analysis have been developed. The statically determinate method assumes pin joints at the mid-points of the verticals and chords of each panel. The method, however, is only suitable for girders with parallel chords of constant stiffness and when the loads are applied at the node points. Various modified moment distribution methods have been developed for the analysis of Vierendeel girders which allow for inclined chords, chords of different stiffness in the panels and member widening at the node positions.

The use of computers offers the most accurate and efficient way  of analysing
Vierendeel girders, particularly those with inclined chords, chords of varying stiffness and when the loading is not applied at the node positions.A further advantage of computer analysis is that joint rotations and deflections are easily calculated.

Plastic theory may be applied to the design of Vierendeel girders in a similar way to its application to other rigid frames such as portal frames. Failure of the structure, as a whole, generally results from local failure of a small number of its members to form a mechanism. Once the failure mode is established the chords and vertical are designed against failure. Computer programs are available for the plastic analysis of plane frameworks including Vierendeel arrangements.

3 Connections
Vierendeel girders have rigid joints with full fixity and so the connections must be of the type which prevents rotation or slip of the incoming members, such as welded or friction-grip bolted connections. Welded connections are usually the most efficient and compact although undesirable if the connections are required to be made on site. Normally site splices are bolted for economy. For very large Vierendeel girders delivered and erected piecemeal, fully bolted connections are normally used.

For member and joint efficiency the ends of the verticals are often splayed. This is of advantage in heavily-loaded girders as the high concentrated local stresses are reduced thus avoiding the need for heavy stiffening.

Some typical joint examples are illustrated in Fig. 19.8(b) and (c).

Fig. 19.8 Typical details of Vierendeel girders: (a) typical forms, (b) welded connections, (c) bolted connections

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Trusses are often used as secondary structures in buildings and bridges in the form of triangulated bracing. Bracing is generally required to resist horizontal loading in buildings or bridges or to prevent deformations and provide torsional rigidity to stiffening girders or box girders.

In buildings, bracing is often required for stability and to transmit horizontal wind loads or crane surges down to foundation level. To avoid the use of heavy com- pression bracing members, the members are usually arranged so that they always act in tension. Although this requires a high degree of redundancy it is normally more economical than providing compression members. Some examples of wind bracing to single-storey building are illustrated in Fig. 19.7(a). For multi-storey buildings with ‘simple’ connections, vertical bracing is required on all elevations to stabilize the building.Normally the floor slabs act as horizontal bracing which transmits the lateral wind loads to the vertical bracing. If the steel frame to the building is erected before the floors are constructed, temporary horizontal bracing must be supplied which can be removed once the floors are in place. Bracing at floor levels may be required if the slabs are discontinuous.Although horizontal bracing in buildings can often be hidden within the depth of the floor, vertical bracing can be obtrusive and undesirable, particularly if the building is clad in glass. In such an instance rigidly-jointed frames are adopted in which the wind loading is resisted by bending in the beams and stanchions. However, if the joints to such frames are made with friction-grip bolts, then temporary bracing may be required for stability prior to the joints being completed. Figure 19.7(b) illustrates some typical bracing systems to multi-storey buildings.

In bridges, secondary truss or bracing frames are often required for stability  and to resist lateral loads due to wind in addition to loads due to road and railway loading such as centrifugal or braking forces. Depending on the  type of structure the bracing may be temporary or permanent, usually placed in both horizontal  and vertical planes. For trusses, permanent plan bracing is provided at both  chord levels for underslung bridges and at deck level for through or semi-through bridges. Where headroom permits, plan bracing is also provided at the top chord level for through trusses, thus conveniently providing restraint to the top com- pression chord and avoiding the need for stiff U-frames. In addition, vertical bracing is also provided between trusses to reduce differential loading and therefore  distortion between trusses and to provide added restraint to the compression chord.

In composite steel plate girder and concrete slab decks temporary vertical bracing may be required when the concrete is poured to provide lateral restraint to the plate girder compression flanges. It may be removed once the concrete has gained sufficient strength to act compositely with the steelwork. In stiffening girders or box girders bracing is often provided in place of plated diaphragms to avoid torsional distortion and to maintain the shape of the cross section under service loads. Figure 19.7(c) illustrates some uses of trusses in bridgeworks. Other uses of trusses in bridgeworks include trestling, i.e. triangulated temporary support frames normally used to support medium- to large-span bridges over land during  erection: see  Fig. 19.7(d).

Fig. 19.7 Other applications of trusses. (a) Bracing to single-storey building; (b) bracing to multi-storey building; (c) typical bracing to bridges; (d) other uses

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Some of the general factors dictating the economy of trusses relating to truss type, spacing, span-to-depth ratios, pitch, etc. have already been discussed earlier.

However, factors such as the location of the structure, contractors’ experience and material availability may have a significant effect on the choice of truss type and details adopted. When designing trusses for overseas locations, particularly the developing countries or for remote areas with difficult access, the designer should consider the following:

• Material available locally, i.e. weldable or unweldable steel
• Preferred connections, i.e. welded, bolted or riveted
• Maximum size of elements that can be transported to site
• Method of erection, capacity and type of plant available
• Use of rolled, compound or fabricated sections
• Redundancy of structure in case of overloading and lack of maintenance
• Experience of local contractors
• Simple design with maximum repetition.

The designer should always try to maximize the use of local materials, labour and expertise so as to avoid expensive importation of materials and trained manpower.

In addition, the relative costs between materials and labour should be reflected in the design.

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1 Design loads
The current British Standards for steel structures in buildings and bridges are both limit-state codes.The magnitude of the partial load factors to be applied is dependent on the load type, the load combination and the limit state (ultimate or serviceability) under consideration.

The following approach may be adopted in deriving the critical  load combinations for each truss member:

(1) The member forces and moments are calculated for each, unfactored, load type (dead, superimposed dead, imposed, wind, etc.) using an appropriate method of analysis.
(2) Load combinations are identified and the appropriate load factors for each combination applied for both serviceability and ultimate limit states.
(3) The critical loads in each element and joint are extracted for both limit states.

The above process is long-winded but with experience the designer can often take short cuts in determining the critical load combinations for each element.

In the analysis the member forces and moments due to joint fixity should be calculated and superimposed on the global member forces. For trusses in buildings the secondary effects due to joint fixity may normally be ignored provided the  slenderness, in the plane of the truss, is greater than 50 for the chord elements and 100 for most of the web members. If this condition is  satisfied the members are assumed to be pin jointed in the analysis. Secondary effects due to axial deformations are usually ignored in building trusses. Local effects due to joint eccentricities and where loads are not applied at nodes should  be taken into account.

For bridge trusses to BS 5400: Part 3: 2000, the effects of joint rigidity are required to be taken into account. Secondary stresses due to axial deformations may be ignored at the ultimate limit state but should be considered at the serviceability limit state and for fatigue checks. As for building trusses, the local effects due to joint eccentricities and cases where loads are not applied at nodes must be considered in bridge trusses.

2 Effective length of compression members
For building trusses the fixity of the joints and the rigidity of adjacent members  may be taken into account for the purpose of calculating the effective length of  compression members. The designer should be careful to ensure that the critical slenderness is identified. For chords, out-of-plane unrestrained lengths do not necessarily relate to the truss nodes, and effective length factors are usually unity; in-plane effective length factors may be demonstrated to be less than unity if the restraining actions of tension members and non-critical compression members are mobilized at the ends of the member. Single angle elements, for both the webs and chord, have minimum radii of gyration that do not lie either in, or normal to, the plane of the truss.

For compression members in bridge trusses the effective lengths may either be obtained from Table 11 of BS 5400: Part 3: 1982 or be determined  by an elastic  critical buckling analysis of the truss.

In the case of simply supported underslung trusses the top compression chord will be effectively restrained laterally throughout its length provided the connection between the chord and the deck is capable of resisting a uniformly distributed lateral force of 2.5% of the maximum force in the chord. The effective length in such a case is taken as zero where friction provides the restraint, or as equal to the spacing of discrete connections where these are provided.

The economic advantages of underslung trusses over through or semi-through trusses is obvious in this respect, due to the dual function of the deck structure.

In the case of unbraced compression chords, that is, chords with  no lateral restraints, the provision of U-frames is necessary. The effective length of the com- pression chord is a function of the stiffness of the chord and  the spacing and stiffness of the U-frame members. Clause 12.5 of BS 5400: Part 3: 1982 gives guidance on the calculation of the effective length of compression chords restrained by U-frames. 19.6.3 Detailed design

For building trusses to BS 5950 the members need only be designed at the ultimate limit state for strength, stability and fatigue where applicable, and at the serviceability limit state for deflection and durability.
Compression members in bridge trusses to BS 5400 are designed at both the ulti- mate and serviceability limit states. Certain compression members, however, are exempt from the serviceability limit state check as defined in clause 12.2.3 of BS 5400: Part 3. Tension members need only be designed at the ultimate limit state.

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Loads are generally assumed to be applied at the intersection point of the members, so that they are principally subjected to direct stresses. To simplify the analysis the weights of the truss members are assumed to be apportioned to the top and bottom chord panel points and the truss members are assumed to be pinned at their ends, even though this is usually not the case. Normally chords are continuous and the connections are either welded or contain multiple bolts; such joints tend to restrict relative rotations of the members at the nodes and end moments develop.

Generally, in light building trusses secondary stresses are negligible and are often ignored. Secondary stresses in light building trusses may be neglected provided that:

• the slenderness of the chord members in the plane of the truss is greater than 50, and
• the slenderness of most of the web members, about the same axis, is greater than 100.

However, in bridge trusses the secondary stresses can be a significant proportion of the primary stresses and must be taken into account. The British Standard for the design of steel bridges, BS 5400: Part 3: 2000, requires the fixity of the joints to be taken into account although axial deformation of the members may be ignored for the ultimate limit state.

The magnitude of the secondary stresses depends on a number of factors including member layout, joint rigidity, the relative stiffness of the incoming members at the joints and lack of fit.

Manual methods of analysis may be used to analyse the stresses, particularly in simple trusses. For simple, statically-determinate trusses,methods of analysis include joint resolution, graphical analysis (Bow’s notation or Maxwell diagram) and the method of sections. The last method is particularly useful as the designer can limit the analysis to the critical sections.

Statically-indeterminate trusses are more laborious to analyse manually;methods available include virtual work, least work and the reciprocal theorem with influence lines. For a full discussion on these methods of analysis the reader should refer to textbooks on structural analysis.

Computers are nowadays readily available to designers and provide a useful means of analysing the most complex of trusses. In addition, joint and member rigidities can easily be incorporated in the modelling thus avoiding laborious hand calculations in determining out-of-balance moments caused by joint deformations.

Local stresses caused by loads not applied at the panel points, joint eccentricities and axial deformation should generally be calculated and superimposed on the direct stresses. However, stresses due to axial deformation are normally neglected except for bridge trusses and trusses of major importance.

Careful consideration must be given to the out-of-plane stability of a truss and resistance to lateral loads such as wind loads or eccentric loads causing torsion about their longitudinal axis.An individual truss is very inefficient, and generally sufficient bracing must be provided between trusses to prevent instability. In bridges, plan bracing is normally provided between trusses at the chord levels in addition to stiff end portals to prevent lateral instability.

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1 Elements
For light roof trusses in buildings the individual members are normally chosen from rolled sections for economy; these are illustrated in Fig. 19.5(a). Structural hollow sections are becoming more popular due to their efficiency in compression and their neat and pleasing appearance in the case of exposed trusses. Structural hollow sections, however, have higher fabrication costs and are only suited to welded construction. For larger-span heavily-loaded roof trusses and small-span bridge trusses it often becomes necessary to use heavier sections such as rolled universal beams and columns and multiples of the smaller rolled sections such as back-to-back angles and channels, Fig. 19.5(b).

Fig. 19.5 Typical element cross sections. (a) Light building trusses; (b) heavy building trusses and light, small-span bridges; (c) road and railway bridges

For large-span bridge trusses, compound or fabricated sections are normally necessary, particularly for the chords and the compression web members. Figure 19.5(c) illustrates some typical arrangements often used, although the choice available to the designer is very wide. For heavily-loaded bridge trusses, fabricated box sections offer economy in material due to their high efficiency in compression.The open sections shown assist the connections to the web members of the truss but require lacing or battening and are prone to distortion during fabrication.

Providing suitable access to all members and surfaces for inspection, cleaning and painting should be a primary consideration in deciding on the sections and details to be incorporated in the design. Laced sections are disadvantageous in this respect as access can be severely restricted. In highly corrosive environments welded closed box or circular hollow sections with welded connections are usually used in order to reduce maintenance costs as all exposed surfaces are readily accessible.

2 Connections
There are basically three types of connections used for connecting truss elements to each other, that is, welding, bolting and riveting.Riveting is rarely used in the UK due to the very high labour costs involved, although it is still widely used in developing countries where labour costs are low. Small-span trusses which can be trans- ported whole from the fabrication shop to the site can be entirely welded. In the case of large-span roof trusses which cannot be transported whole, welded sub- components are delivered to site and are either bolted or welded together on site.

Generally in steelwork construction bolted site splices are much preferred to welded splices for economy and speed of erection. In light-building roof trusses entirely bolted connections are less favoured than welded connections due to the increased fabrication costs, and usually bolted connections require cumbersome and obtru- sive gusset plates. However, bolted connections are more widely used in bridge trusses, particularly medium- to large-span road bridges and railway bridges, due to their improved performance under fatigue loading. In addition, bolted connections may sometimes permit site erection of the individual elements without the need for expensive heavy craneage. Gusset plates are often associated with bolted bridge trusses, their size being dependent on the size of the incoming members and the space available for bolting.

Gusset plates also enable the incoming members to be positioned in such a way that their centroidal axes meet at a single point, thus avoiding load eccentricities.

Ideally for all types of trusses the connections should be arranged so that the centres of gravity of all incoming members meeting at the joint coincide. If this is not possible the out-of-balance moments caused by the eccentricities must be taken into account in the design.

Some typical joint details are illustrated in Fig. 19.6.

Fig. 19.6 Typical joints in trusses. (a) Welded RHS building roof truss; (b) bolted bridge truss